Written by Brandon Chi, Engineering Manager, Lateral Systems at Simpson Strong-Tie.
Wood shearwalls are typically used as a lateral-force-resisting system to counter the effects of lateral loads. Wood shearwalls need to be designed for shear forces (using sheathing and nailing), overturning (using holdowns), sliding (using anchorage to concrete) and drift, to list some of the main dangers. The Simpson Site-Built Shearwall Designer (SBSD) web app is a quick and easy tool to design a wood shearwall based on demand load, wall geometry and design parameters.
The web application provides two options for generating an engineered shearwall solution: (1) Solid Walls; and (2) Walls with Opening using the force-transfer-around opening (FTAO) method. Both options generate solutions that offer different combinations of sheathing, nailing, holdowns, end studs and number/type of shear anchors. The app can generate a PDF output for each of the possible solutions. Design files can be saved and reused for future projects.
Figure 1 shows the input screens for the “Solid Walls” and “Walls with Opening” designs with common wall parameters that are applicable to both design options. The user interface uses quick drop-down menu and input fields for the designer to select the different options and parameters. Unless otherwise noted, all the input loads are to be nominal (un-factored) design loads. The application will apply load combinations to determine the maximum demand forces for the shearwall design.
Figure 2 shows the allowable stress design (ASD) load combinations used for calculating the demand loads for the different components of the wood shearwall (i.e., holdown, compression post, sheathing and nailing design, etc.).
In addition to the lateral loads (wind and seismic) applied at the top of the wall and the wall’s own weight, uniform loads on top of the wall and concentrated point loads at the end posts can also be modeled. (See Figure 3.)
Embedded anchor or embedded strap holdowns can be modeled by the app. (See Figure 4.) For the embedded strap option, additional input parameters are required since they will affect the allowable load of the selected strap holdown.
The Designer has the option to include additional sources of vertical displacement for drift calculation. (See Figure 5.)
For hand-calculated design when the demand forces are determined, the holdown size and shear anchorage can be selected from tabulated values. Design for the sheathing/nailing and compression post is relatively straightforward as well; however, the shearwall drift calculation may take a bit more work. This is where the SBSD app comes in handy. Below are two sections on the shearwall drift and strap force calculations and assumptions used in the SBSD application. If you are interested, please contact Simpson Strong-Tie for other design assumptions used in designing the SBSD app.
The Δa value from the third term of the equation is the total vertical elongation of the wall holdown system from the applied shear in the shearwall. The third term accounts for the additional displacement from holdown displacement. For holdown deflection, the deflection value depends on the post size used with the holdown size. When hand-calculating shearwall drift, Designers may have to perform a couple of iterations to come to the final post and holdown size. The SBSD app accounts for the holdown displacement and the post size used for overturning force calculation.
For shearwall-with-opening deflection calculation, EQ-2 is used in the SBSD app.
The solid wall, ∆solid wall, term is calculated using EQ-1 above. For the window strip and wall pier deflection terms, the height “h” used in EQ-1 is taken as the height of the window opening. ∆a is the deflection from nail slip in the shearwall. For more information regarding shearwall deflection with opening, please refer to Example 1 in Volume 2 of the 2015 IBC SEAOC Structural/Seismic Design Manual.
Strap Force Calculations:
For the Wall with Opening design option, there are several methods (Drag Strut, Cantilever Beam, SEAOC/Tompson, Diekmann) to calculate the force transfer around the opening. In the SBSD app, the Diekmann technique is used to calculate the pier forces in the shearwall and the strap forces around the opening. When calculating the strap forces, the SBSD app assumes they are the same at the top and bottom of the opening. In addition, contribution of the gravity load only affects the overturning forces in the holdown and post design but not the wall pier forces or strap forces.
Once all design parameters are entered and calculated, a list of possible solutions (where available) will be shown. (See Figure 6.) Common parameters such as sheathing material and type, wood species, minimum lumber grade, etc., are shown first, followed by other design parameters. The user can filter the solutions by seismic drift or wind drift.
The Designer can select the PDF button next to the desired solution to see a PDF design file on a separate screen. (See Figure 7.) The PDF design file contains the detailed design criteria input by the Designer, calculated demand loads, shearwall material summary, and a design summary for holdown, sheathing, and compression post design. A detail summary for shearwall deflection is also shown, with each term of the shearwall deflection equation (EQ-1) separated. Shear anchorage and design assumption notes follow the design summary section. This PDF file can be saved and printed by the Designer.
I hope you find the SBSD web app helpful for your day-to-day wood shearwall design needs. If you have any questions or comments, please leave them in the comments section below.
How would a six-story light-frame wood building perform in a large earthquake? Back in 2009, Simpson Strong-Tie was a partner in the World’s Largest Earthquake Test, a collaboration of the NEESWood project, to answer that question. This was a full-scale test which subjected the building to 180% of the Northridge earthquake ground motions (approximately a M7.5). Within the building, Simpson Strong-Tie connectors and Strong-Frame SMF were used, with the Strong-Rod™ anchor tiedown system (ATS) serving as holdown for each shearwall.
The NEESWood building was designed under Performance-Based Design methodology, and the test was conducted as validation for the approach. Buildings of similar size to the NEESWood building are built to current codes using similar products. Mid-rise light-frame wood structures continue to be a popular form of construction in various densely populated cities across the country. As part of the lateral-force-resisting system, continuous rod systems are used as the holdown for the shearwall overturning restraints. Simpson Strong-Tie has been involved with continuous rod systems since the early 2000s when we launched the Strong-Rod anchor tiedown system.
Today, rod manufacturers design the continuous rod systems with design requirements (loading, geometry, etc.) Supporting documents (e.g., installation details, layouts, RFI/markups and calculations) are submitted for each unique project. Over the years, engineers have asked many questions related to the design of these systems. In this week’s blog, we will explore Frequently Asked Questions pertaining to Strong-Rod ATS systems used as shearwall overturning restraints (holdowns).
The majority of these components are designed in accordance with the building code and reference standards (e.g., NDS, AISC). A project-specific calculation package is submitted for each job that addresses the evaluation of these elements. Therefore, these elements are not listed in evaluation reports.
Shrinkage compensation devices, on the other hand, are proprietary components which are not addressed by the building code or reference standards. Therefore, they are tested in accordance with ICC-ES acceptance criteria AC316 and are listed in ICC-ES ESR-2320.
What is the material specification of the rods used above concrete?
The specified rod materials are shown in Table 1.
Can threaded rods or couplers be welded to steel beams?
Simpson Strong-Tie generally does not recommend this practice. Of the materials listed in Table 1, ASTM A307 material is the only specification that contains supplementary requirements for welding. When standard strength rod is supplied to the job, it is not guaranteed that this will be the material provided.
ASTM A449 and A193-B7 high-strength rods develop strength and ductility characteristics through controlled quenching and tempering treatments. Quenching is the rapid cooling of metal (usually by water or oil) to increase toughness and strength. This process often increases brittleness. Tempering is a controlled reheating of the metal which increases ductility after the quenching process. Precise timing in the application of temperature during the tempering process is critical to achieving a material with well-balanced mechanical properties. It is unlikely that field welding will satisfy the requirements of quenching and tempering.
Coupler nuts are generally fabricated from material exhibiting characteristics similar to high-strength rods. Thus, it is not recommended to weld coupler nuts to steel beams due to the potential for embrittlement.
Simpson Strong-Tie specifies a weldable cage which is fabricated from ASTM A36 material for such applications.
How do you calculate the Maximum ASD Tension Capacity provided in the job summary?
Simpson Strong-Tie provides a comprehensive design package for continuous rod systems used as holdowns for multi-story stacked shearwalls. The individual run calculations, as shown in Figure 1, provide the Maximum Tension Capacity, which correlates to the maximum force the system can deliver. Plan check often requests justification on how these values are derived at each level. These values are calculated, and the process explained below may be used on any Simpson Strong-Tie ATS Job Summary as justification.
The maximum tension capacity published within the Job Summary and the Installation Details is derived using the following procedure:
Step 1: Evaluate the top-most level. Compare the published capacities of the rod in tension, plate in bearing and the take-up device. The lowest of these three will govern and becomes the Maximum Tension Capacity for this level.
Step 2: Evaluate the next level down. (a) Sum the Maximum Tension Capacity from Step 1 and the published capacity of the take-up device from this level. (b) Sum the Maximum Tension Capacity from Step 1 and the published capacity of the plate in bearing from this level. (c) Compare derived values from (a) and (b) to the published capacity of rod in tension. The lowest of these three values will govern and becomes the Maximum Tension Capacity of this level.
Step 3: Repeat Step 2 as necessary until the bottom-most level is reached.
Applying this procedure to the sample run, SW9, will wield the following result:
Step 1: Evaluate capacities published at Level 4
Plate in bearing (PBRTUD5-6A) = 7.06 kipsgoverns
Take-up device (RTUD6) = 20.83 kips
Rod in tension (ATS-R6) = 9.61 kips
The lowest value in Step 1 is the plate in bearing, hence 7.06 kips is the maximum load that can be delivered at Level 4 and is the Maximum Tension Capacity.
Step 2: Evaluate capacities at Level 3
Maximum Tension Capacity from Level 4 = 7.06 kips (See Step 1)
Maximum Tension Capacity from Level 4 + take-up device (ATS-ATUD9-2) = 7.06 + 12.79 = 19.85 kips
Maximum Tension Capacity from Level 4 + plate in bearing (PL9-3×5.5) = 7.06 + 10.03 = 17.09 kips
Rod in tension (ATS-R7) = 13.08 kipsgoverns
The lowest value in Step 2 is the rod in tension, hence 13.08 kips is the maximum load that can be delivered at Level 3 and is the Maximum Tension Capacity.
Step 3: Evaluate capacities at Level 2
Maximum Tension Capacity from Level 3 = 13.08 kips (See Step 2)
Maximum Tension Capacity from Level 3 + take-up device (ATS-ATUD9-2) = 13.08 + 15.56 = 28.64 kips
Maximum Tension Capacity from Level 3 + plate in bearing (PL9-3×5.5) = 13.08 + 10.03 = 23.11 kips
Rod in tension (ATS-R7) = 13.08 kipsgoverns
The lowest value in Step 3 is the rod in tension, hence 13.08 kips is the maximum load that can be delivered at Level 2 and is the Maximum Tension Capacity.
Step 4: Evaluate capacities at Level 1
Maximum Tension Capacity from Level 2 = 13.08 kips (See Step 3)
Maximum Tension Capacity from Level 2 + take-up device (ATS-ATUD14) = 13.08 + 24.39 = 37.47 kips
Maximum Tension Capacity from Level 2 + plate in bearing (PL14-3×8.5) = 13.08 + 13.98 = 27.05 kipsgoverns
Rod in tension (ATS-R11) = 32.30 kips
The lowest value in Step 4 is due to the plate in bearing, hence 27.05 kips is the maximum load that can be delivered at Level 1 and is the Maximum Tension Capacity.
In the System Deflection Summary page(s) of the Job Summary, is the Total System Deflection provided at Allowable or Strength levels?
Immediately following the individual run calculations in each job summary, Simpson Strong-Tie provides a summary of deflection of the rod system similar to what is shown in Figure 2. This breaks down the deformation of all components being considered. In the example below, the rod elongation and deflection of the take-up device are summed to provide the total deflection.
The calculated system deflection is presented at ASD level. See section below for how to use these system deflections for your drift calculation.
What system deflection limit do you typically design to, and what does that include?
Unless otherwise specified on the plans or required by the building jurisdiction, Simpson Strong-Tie will design the continuous rod system to satisfy the deformation limits set forth in ICC-ES Acceptance Criteria (AC316). In some instances, the Designer may need a more restrictive deformation due to project specific conditions (e.g., tight building separations) and will require rod manufacturers to design for a lower deformation. Some jurisdictions (e.g., City of San Diego, City of San Francisco) may also have specific design requirements that continuous rod systems must conform to. The minimum recommended per-floor deformation limit set forth in AC316 is:
PD = ASD demand cumulative tension load (kips) L = length of the rod between restraints – i.e., floor-to-floor (in.) A = net tensile area of the rod (in.2) E = Young’s Modulus of Elasticity (29,000 ksi) ΔR = seating increment of the shrinkage compensation device (as published in ICC-ES evaluation report) ΔA = deflection of the shrinkage compensation device at the allowable load (as published in ICC-ES evaluation report) PA = Allowable capacity (kips)
Should deformation limits be specified in the construction documents?
Simpson Strong-Tie strongly recommends this information be included in the construction documents. Along with the cumulative tension and compression forces, the required deformation limits for the holdown are important to ensure that rod manufacturers are designing the holdown to satisfy the desired shearwall performance.
How do I use the system deformation limit?
The System Deflection is the total deformation of the holdown system from floor to floor (refer to the last two columns in Figure 2). This information represents the total ASD holdown deformation term, Δa, for each level and is to be used in the shearwall drift equation from the Special Design Provisions for Wind and Seismic (2015 SDPWS 4.3-1).
ASCE 12.8.6 requires that shearwall drift be calculated at strength level. Therefore, the information provided within the System Deflection Summary page needs to be converted from ASD to Strength Level. The conversion factors in Table 2 can be used to convert the ASD deformations to strength level. For discussions and methodology in converting bearing plate deformation to strength level, please refer to the WoodWorks Design Example of a Five-Story Wood Frame Structure over Podium Slab found here.
Can rod systems be used in Type III construction?
Yes! 2015 IBC §2303.2.5 requires that Fire Retardant-Treated Wood (FRTW) design values be adjusted based on the type of treatment used on the project. Adjustment factors vary for each FRTW manufacturer; refer to the ICC-ES evaluation report of the specified FRTW manufacturer for the unique adjustment values. Rod manufacturers need to know what treatment is being used so this information can be taken into consideration when designing compression posts and incremental bearing (bearing plates).
What are Simpson Strong-Tie’s guidelines for fire caulking material?
While there are many options for fire-rated caulking, these products can be used in conjunction with the Simpson Strong-Tie ATS system. Below is a list of considerations when selecting and specifying a material for use where the rods penetrate the top and sole plates:
The fire-rated caulking shall not be corrosive to metal when used in contact with ATS components.
Direct contact with shrinkage compensating devices (e.g., TUD, ATUD, RTUD) shall be avoided. Shrinkage compensating devices have moving components and may not function properly with debris interference.
Indirect contact with shrinkage compensating devices shall also be avoided. Shrinkage compensation accumulates up the building and therefore the largest shrinkage occurs at the top of the building. As such, when the building shrinks, remnants of the material may still be stuck to the threads of the rod and may be detrimental to the performance of some shrinkage compensating devices (e.g., an RTUD). It is recommended to detail the installation with shrinkage taken into consideration.
The fire-rated caulking should be pliable to accommodate wood shrinkage and the building moving down during this process.
The performance and the suitability of fire-rated caulking are outside the scope of Simpson Strong-Tie.
Why doesn’t your design include compression post design?
If the Engineer of Record has already specified compression posts to be used with a continuous rod system, Simpson Strong-Tie will not provide these on the holdown installation drawings. This is primarily done to prevent discrepancies between the specification in the contract documents and what is shown on the installation drawings.
What is the maximum spacing between compression posts?
For platform-framed structures, the maximum spacing between compression posts is 9″. The large majority of Simpson Strong-Tie bearing plates will fit within the 9″ spacing requirement, eliminating the need for notching compression posts. In some framing conditions, such as balloon framing or a top chord bearing truss, the maximum spacing will be reduced to 6″. This is due to the limited amount of space between the top of the compression posts transferring uplift (via bearing) into the point of restraint (e.g., bearing plate) at the level above. To ensure this load path is complete, the posts need to be spaced closer.
What is the nailing schedule for the bridge block to the king studs?
Simpson Strong-Tie doesn’t recommend nailing the bridge block to the cripple as the bridge block member will shrink. Locking the bridge block in place may result in a gap forming between the bottom of the bridge block member and the top of the cripple studs, which is not accounted for in the Total System Deflection.
Are there any published documents with design examples of continuous rod systems used in mid-rise construction?
This blog post continues our series on the final results of the 2016 ICC Group B Code Change Hearings. This post will focus on approved changes to the International Residential Code (IRC) that are of a structural nature. The changes outlined here will be contained in the 2018 IRC, which is expected to be published in the fall of this year.
In Chapter 3, the seismic design category / short-period design spectral response acceleration maps will be updated to match the new USGS/NEHRP Seismic Maps. These new maps are based on the worst case assumption for Site Class. Significantly, a new set of maps will be provided in Figure 301.2(3) “Alternate Seismic Design Categories”. These are permitted to be used when the “soil conditions are determined by the building official to be Site Class A, B or D.” See page 29 of the linked document for the new maps and a good explanation of the changes that will be occurring in various parts of the country. In addition, the ICC Building Code Action Committee authored a reorganization of the seismic provisions of Chapter 3 to try to reduce confusion.
Another change in Chapter 3 will clarify that guards are only required on those portions of walking surfaces that are located more than 30 inches above grade, not along the entire surface. To bring consistency with the IBC, another change will require that staples in treated wood be made of stainless steel.
A broad group of parties interested in deck safety, known as the Deck Code Coalition, submitted 17 different code changes with revisions to Section R507 on decks. Of those, 12 were approved, making significant changes to that section. The various approved changes included the following: a complete re-write of that section; new/clarified requirements for deck materials, including wood, fasteners and connectors; clarified requirements for vertical and lateral connections of the deck to the supporting structure; new requirements for sizes of deck footings and specification that deck footings must extend below the frost line, with certain exceptions; clarification for deck board material, including an allowance for alternative decking materials and fastening methods; adding new columns to the deck joist span table that show the maximum cantilever for joists; adding the allowance for 8×8 deck posts, to allow notching for the support of a three-ply beam; and clarification of the deck-post-to-footing connection.
In Chapter 6, a new table permitting 11ʹ- and 12ʹ- long studs was added. In the 2015 IRC, load-bearing studs were limited to 10 feet in length. A new high-capacity nail, the RSRS (Roof Sheathing Ring Shank) nail was added as an option for fastening roof sheathing. This nail will become more widely used once the higher roof component and cladding forces from ASCE 7-16 are adopted. The rim board header detail that was added for the 2015 IRC was corrected to show that hangers are required in all cases when the joists occur over the wall opening.
There were several changes made to the Wall Bracing Section, R602.10. The use of the 2.0 increase factor was clarified for use when the horizontal joints in braced wall panel sheathing are not blocked. Narrow methods were added to the column headings for the wind and seismic bracing amount tables, to make them consistent, and the methods for adding different bracing were clarified. When using bracing method PFH, the builder can omit the nailing of the sheathing to the framing behind the strap-type holdown. Finally, offering some relief for high-seismic areas where brick veneer is used, an allowance was added to permit a limited amount of brick veneer to be present on the second floor without triggering the use of the BV-WSP bracing method.
In Chapter 8, the requirements for a “stick-framed” roof system were completely re-written to make such systems easier to use.
A couple of significant changes were made to the prescriptive requirements for cold-formed steel framing. The requirements for the anchorage of cold-formed steel walls were revised, and the wind requirements for cold-formed steel framing were changed to match the new AISI S230 prescriptive standard.
Finally, it may be helpful to mention some of the proposed changes that were not adopted. While the new ASCE 7-16 was adopted as the IRC reference standard for loads as part of the Administrative changes, several changes to the IRC to make it consistent with ASCE 7-16 were not approved. A change to update the IRC wind speed maps, roof component and cladding pressures, component and cladding roof zones, and revise the remainder of wind-based requirements to match ASCE 7-16 was not approved. Similarly, a proposal to increase the live load on decks, from 1.0 to 1.5 times the occupancy served, was denied.
Once the IRC is published, it will be time to start a new code change cycle once again, with Group A code changes due January 8, 2018. The schedule for the next cycle is already posted here.
What changes would you like to see for the 2021 codes?
In a Structural Engineering Blog post I wrote last October, “Soft-Story Retrofits Using the New Simpson Strong-Tie Retrofit Design Guide,” one item I barely touched on at the time was the benefit of using Simpson Strong-Tie® Strong Frame special moment frames to retrofit vulnerable soft-story wood-framed buildings commonly found on the West Coast. In this post, I will be diving into more detail on a few features that make the Strong Frame special moment frame truly special.
In the recent release of the ANSI/AISC 358-16 (AISC 358-16), the Simpson Strong-Tie Strong Frame moment connection has been included as a prequalified special moment frame (SMF) connection. Prequalified moment connections are structural-steel moment connection configurations and details that have been reviewed by the AISC Connection Prequalification Review Panel (CPRP) and incorporated into the AISC 358 standard. What’s unique about this newly prequalified connection is that it’s the first moment connection to be prequalified in AISC as a partially restrained (PR-Type) moment connection.
With this recent inclusion into AISC 358-16, we’ve also developed our newly released Strong Frame Design Guide to help designers understand the differences in design and detailing between the Strong Frame connection and traditional SMF connections. The following are just a few of the key differences discussed in this guide.
SMF Yielding Elements
Traditional prequalified moment frames most often require a welded connection with either a weakened beam or a stiffened connection. SMF connections are designed so that the beam will yield as necessary under large displacements that may occur during a seismic event. The yielding of the beam section provides energy dissipation and is designed to ensure that the fully restrained beam-to-column connection isn’t compromised. The current design philosophy is the product of extensive testing of SMF connections based on studying the effects of the 1994 Northridge and 1989 Loma Prieta earthquakes in California. Figures 1, 2 and 3 below depict test specimens that demonstrate yielding at the designated areas of the beam.
The Strong Frame SMF has taken a different approach to the traditional connections by utilizing a Yield-Link® structural fuse designed to provide the energy dissipation for the beam-to-column moment connection. This is a modified T-Stub that has a reduced section in the stem. The yielding during a seismic event has been moved from the beams to the Yield-Link structural fuse. The fuse can be replaced after a major event, very much like an electrical fuse when overloaded. A traditional moment frame may require a much more invasive structural repair.
Beam Lateral Bracing
The traditional types of prequalified connections, as along with other proprietary connections included in AISC 358, all require the beam to yield so as to dissipate energy as discussed above. These types of connections require that the beam be braced to resist the lateral torsional buckling per code. However, it is difficult to meet the bracing requirements in the case of a steel SMF in a wood structure.
With the Strong Frame SMF connection, the energy dissipation is moved from the beams to the Yield-Link structural fuses, with the connection following a capacity-based design approach. This allows the connection to remain elastic under factored load combinations. With the yielding confined to the structural fuses, inelastic deformation is not expected from the members and lateral beam buckling braces are not required. The beam can be designed to span the entire length without beam bracing. See also this blog post.
Column-Beam Relationship Requirements
Traditional SMF follow a strong column – weak beam requirement to ensure plastic hinging occurs in the beams and not the columns. If the energy dissipation takes place within such hinging in the beams, the column members will remain elastic so as to provide stability and strength for the above stories. If plastic hinges occur in the columns, there is a potential for the formation of a weak-story mechanism.
The Strong Frame special moment frame is unlike the traditional SMF, where the plastic hinges are formed by the buckling of the beam flange and web. In the Strong Frame SMF, the stretching and shortening of the links at the top and bottom of the Strong Frame beams are the yielding mechanisms. So instead of a strong column – weak beam check, the Strong Frame design procedure checks for a strong column – weak link condition where the ratio of the column moments to the moment created by the Yield-Link® couple is required to be greater than or equal to 1.0.
Traditional moment frame connections typically require welding in the field. Where bolted SMF connections are used, pretensioned bolts are necessary. Both welding and pretensioned bolts require third-party special inspection.
The Strong Frame SMF has been designed and tested as a 100% field-bolted connection. Unlike other bolted options, the Strong Frame’s field-bolted connections only need to be made snug tight. No onsite bolt pretensioning or special inspections are required with this system. This allows the beams and columns to be maneuvered into place, erected and installed in a fraction of the time needed for the welding, lateral-beam-bracing installation and additional inspections or repairs that traditional moment frames typically require.
One last item I’d like to discuss is the design service that Simpson Strong-Tie provides for the Strong Frame special moment frame. Whether you design moment frames only once in a while or on a regular basis, the Strong Frame design team will provide you with No-Equal design support at no additional cost. Designers receive a complete package that includes drawings and calculations, which are submittal-ready. This ensures that you’ll have a frame connection design meeting the latest codes and design requirements. Contact email@example.com for more information or to request design support.
To learn more about the special benefits and uses of Strong Frame moment frames, check out the following links:
One of the ways I get through winter every year is by looking forward to the weekend in March when we set our clocks ahead and “spring forward” into Daylight Savings time. Some people don’t like this change because of the lost hour of sleep, but to me it means the weather shouldn’t be cold for much longer.
The coming of spring means getting to walk to the car in daylight at the end of the workday. It also means getting the garden started for the year and spending more time outside in general.
Of course, I’m not alone in being happy to see winter go.
In the residential world, the phenomenon of “deck season” coincides with this time of year. Homeowners with decks are getting ready for summer by giving their decks a cleaning and looking them over for any needed maintenance. Now’s the time that new or replacement decks are being planned and built to be enjoyed for the rest of the year.
It’s no coincidence, then, that our deck-code guide has been updated again in time for warmer weather. The Deck Connection and Fastening Guide goes detail by detail (ledger connection, joist-to-beam connection, beam-to-post connection, etc.) through a typical deck and identifies the relevant building-code requirements (2012 and 2015 IRC/IBC) and connection options.
Our deck-code guide can be a helpful reference to an engineer who is just getting acquainted with decks, and can also bring you up to speed on revisions to the IRC that can necessitate engineering changes to even a relatively simple residential deck. Multilevel decks, guardrail details, ledger details and foundation challenges are all examples of things a deck builder could call you for assistance with.
Last year, I gave a presentation at the annual National Council of Structural Engineers Associations (NCSEA) Summit in Orlando, Florida, titled “Becoming a Trusted Advisor: Communication and Selling Skills for Structural Engineers.” As this was a summit for the leaders of the structural engineers associations from across the country, I wasn’t sure how many people would find it valuable to spend their time learning about a very nontechnical topic. To my surprise and delight, the seminar ended up being standing-room only, and I was able to field some great questions from the audience about how they could improve their selling and communication skills. In the many conversations I had with the conference attendees after my presentation, the common theme was that engineers felt they needed more soft-skills training in order to better serve their clients. The problem, however, was finding the time to do so when faced with the daily grind of design work.
When I started my first job as a design engineer at a structural engineering consulting firm straight out of school, I was very focused on improving and expanding my technical expertise. Whenever possible, I would attend building-code seminars, design reviews and new product solution presentations, all in an effort to learn more about structural engineering. What I found as I progressed through my career, however, was that no matter how much I learned or how hardworking I was, it didn’t really matter if I couldn’t successfully convey my knowledge or ideas to the person who really mattered most: the client.
How can an engineer be most effective in explaining a proposed action or solution to a client? You have to be able to effectively sell your idea by understanding the needs of your client as well as any reasons for hesitation. The importance of effective communication and persuasion is probably intuitive to anyone who’s been on the sales side of the business, but not something that occurs naturally to data-driven folks like engineers. As a result of recent legislation in California, however, structural engineers are starting to be inundated with questions from a group of folks who have suddenly found themselves responsible for seismically upgrading their properties: apartment building owners in San Francisco and Los Angeles.
Imagine for a moment that you are a building owner who has received a soft-story retrofit notice under the City of Los Angeles’ Ordinance 183893; you have zero knowledge of structural engineering or what this term “soft-story” even means. Who will be your trusted advisor to help you sort it out? The City of Los Angeles Department of Building and Safety (LADBS) has put together a helpful mandatory ordinance website that explains the programs and also offers an FAQ for building owners that lets them know the first step in the process: hire an engineer or architect licensed in the state of California to evaluate the building.
I’ve had the opportunity to be the first point of contact for a building owner after they received a mandatory notice, because it turns out some relatives own an apartment building with soft-story tuck-under parking. Panicked by the notice, they called me looking to understand why they were being forced to retrofit a building that “never had any problems in the past.” They were worried they would lose rent money due to tenants needing to relocate, worried about how to meet the requirements of the ordinance and, most importantly, worried about how much it was going to cost them. What they really wanted was a simple, straightforward answer to their questions, and I did my best to explain the necessity behind retrofitting these vulnerable buildings and give an estimated time frame and cost that I had learned from attending the first Los Angeles Retrofit Resource Fair in April 2016. With close to 18,000 buildings in the cities of San Francisco and Los Angeles alone that have been classified as “soft-story,” this equates to quite a number of building owners who will have similar questions and be searching for answers.
Submitting Building Plans with the Right Retrofit Product Solutions
Communicating with Your Building Tenants
Completing Your Soft-Story Retrofit
We encourage you to invite any clients or potential clients to attend this informative webinar, which will lay the foundation for great communication between the two of you. As part of the webinar, we will be asking the building owners for their comments, questions and feedback so we can better understand what information they need to make informed decisions, and we will be sure to share these with the structural engineering community in a future post. By working together to support better communication and understanding among all stakeholders in retrofit projects, we will be well on our way to creating stronger and more resilient communities!
For additional information or articles of interest, there are several resources available:
In early December, ICC posted the preliminary results of the Group B Online Governmental Consensus Vote, which included structural changes to the IBC, IEBC and IRC. ICC reports that there were more than 162,000 votes cast by eligible Voting Members during the three-week online voting period.
One subject of interest to building Designers, builders and some building-material suppliers was the disposition of a group of code changes that adopted ASCE 7-16 as the reference standard on loads for the IBC and IRC, and changed other parts of the IBC and IRC to reflect that.
The most controversial part of adopting the new ASCE 7-16 standard was its increase in roof component and cladding loads. The higher pressure coefficients in some cases raised the concern that the cost of roofing, roofing materials and roof repairs would be increased. Other items that raised some opposition were the new chapter on tsunami loads and the increase in deck and balcony live loads from 40 psf to 60 psf.
Despite these concerns, ICC members voted to approve the code change that adopted ASCE 7-16 as the reference for loads in the 2018 IBC, IRC and IEBC.
Along with that specific change, several other related changes were approved to correlate the IBC with adoption of ASCE 7-16. These included changes to Section 1604, General Design Requirements; adding in a new Section 1615 on Tsunami Design Requirements; modifications to Section 1613 so that seismic design requirements match ASCE 7-16; and deletion of Section 1609.6, Alternate All-Heights Method for wind design. On this last item, the argument was that since ASCE 7 now includes a simplified wind load design method, a competing method is not needed in the IBC.
Interestingly, a change to remove Strength Design and Allowable Stress Design load combinations from the IBC, which was approved by the IBC Structural Committee, was overturned and denied by the ICC Member voters. So those will remain in the IBC.
For the IRC, even though ASCE 7-16 will be shown as the referenced load standard, most changes to the actual code language relating to the new standard were denied. Items that were specifically denied included adoption of ASCE 7-16 wind speed maps, adoption of ASCE 7-16 roof pressure loading, and adoption of the new higher deck and balcony live loads. So the result is that the IBC and IRC will again be inconsistent with each other regarding wind design. On the other hand, the new USGS/NEHRP Seismic Design Maps were approved.
Future Code Corner articles will address other changes approved for the 2018 IBC and IRC.
As published in STRUCTURE magazine, September 2016. Written by Randy Daudet, P.E., S.E., Product Manager at Simpson Strong-Tie. Re-posted with permission.
One of the world’s greatest unsolved mysteries of our time lies in a courtyard outside of the Central Intelligence Agency (CIA) headquarters in Langley, Virginia. It’s a sculpture called Kryptos, and although it’s been partially solved, it contains an inscription that has puzzled the most renowned cryptanalysts since being erected in 1990. Meanwhile, in another part of the DC Beltway about 15 miles to the southeast, another great mystery is being deciphered at the American and Iron Institute (AISI) headquarters. The mystery, structural behavior of cold-formed steel (CFS) clip angles, has puzzled engineers since the great George Winter helped AISI publish its first Specification in 1946. In particular, engineers have struggled with how thin-plate buckling behavior influences CFS clip angle strength under shear and compression loads. Additionally, there has been considerable debate within the AISI Specification Committee concerning anchor pull-over strength of CFS clip angles subject to tension.
The primary problem has been the lack of test data to explain clip angle structural behavior. Even with modern Finite Element Analysis (FEA) tools, without test data to help establish initial deformations and boundary conditions, FEA models have proven inaccurate. Fortunately, joint funding provided by AISI, the Steel Framing Industry Association (SFIA), and the Steel Stud Manufactures Association (SSMA) has provided the much-needed testing that has culminated in AISI Research Report RP15-2, Load Bearing Clip Angle Design, that summarizes phase one of a multi-year research study. The report summarizes the structural behavior and preliminary design provisions for CFS load bearing clip angles and is based on testing that was carried out in 2014 and 2015 under the direction of Cheng Yu, Ph.D. at the University of North Texas. Yu’s team performed 33 tests for shear, 36 tests for compression, and 38 tests for pull-over due to tension. Clip angles ranged in thickness from 33 mils (20 ga.) to 97 mils (12 ga.), with leg dimensions that are common to the CFS framing industry. All of the test set-ups were designed so that clip angle failure would preclude fastener failure.
For shear, it was found that clips with smaller aspect ratios (L/B < 0.8) failed due to local buckling, while clips with larger aspect ratios failed due to lateral-torsional buckling. Shear test results were compared to the AISC Design Manual for coped beam flanges, but no correlation was found. Instead, a solution based on the Direct Strength Method (DSM) was employed that utilized FEA to develop a buckling coefficient for the standard critical elastic plate-buckling equation. Simplified methods were also developed to limit shear deformations to 1/8 inch. For compression, it was found that flexural buckling was the primary failure mode. Test results were compared to the gusset plate design provisions of AISI S214, North American Standard for Cold-Formed Steel Framing – Truss Design, and the axial compression member design provisions and web crippling design provisions of AISI S100, North American Specification for the Design of Cold-Formed Steel Structural Members, but no good agreement was found. Therefore, an alternate solution was developed that utilized column theory in conjunction with a Whitmore Section approach that yielded good agreement with test results. It was further found that using a buckling coefficient of 0.9 in the critical elastic buckling stress equation will produce conservative results. Finally, for pull-over due to tension, it was found that clip angle specimens exhibited significant deformation before pulling over the fastener heads (essentially the clip turns into a strap before pull-over occurs). However, regardless of this behavior, tested pull-over strength results were essentially half of AISI S100 pull-over equation E4.4.2-1.
Thanks to AISI Research Report RP15-2, there is a clearer understanding of the CFS clip angle structural behavior mysteries that have puzzled engineers for many years. However, just as the CIA’s Kryptos remains only partially solved, some aspects of clip angle behavior remain a mystery. For instance, how are the test results influenced by the fastener pattern? All of the test data to date has used a single line of symmetrically placed screws. This is something that does not occur for many practical CFS framing situations and will need additional research. Another glaring research hole is the load versus deflection behavior of clip angles under tension. As briefly mentioned above, the existing pull-over testing has demonstrated that excessive deflections can be expected before pull-over actually occurs. Obviously, most practical situations will dictate a deflection limit of something like 1/8 inch or 1/4 inch, but today we don’t have the test data to develop a solution. Fortunately, AISI in conjunction with its CFS industry partners continues to fund research on CFS clip angle behavior that will answer these questions, and possibly many more.
Jhalak Vasavada is currently a Research & Development Engineer for Simpson Strong-Tie. She has a bachelor’s degree in civil engineering from Maharaja Sayajirao (M.S.) University of Baroda, Gujarat, India, and a master’s degree in structural engineering from Illinois Institute of Technology, Chicago, IL. After graduation, she worked for an environmental consulting firm called TriHydro Corporation and as a structural engineer with Sargent & Lundy, LLC, based in Chicago, IL. She worked on the design of power plant structures such as chimney foundations, boiler building and turbine building steel design and design of flue gas ductwork. She is a registered Professional Engineer in the State of Michigan.
At Simpson Strong-Tie, we strive to make an engineer’s life easier by developing products that help with design efficiency. Our products are designed and tested to the highest standards, and that gives structural engineers the confidence that they’re using the best product for their application.
Having worked in the design industry for almost a decade, I can attest that having a catalog where you can select a product that solves an engineer’s design dilemma can be a huge time- and money-saving tool. Design engineers are always trying to create efficient designs, although cost and schedule are always constraints. Moment connections can be very efficient — provided they are designed and detailed correctly. With that in mind, we developed a moment post base connector that can resist moment in addition to download, uplift and lateral loads. In this post, I would like to talk about moment-resisting/fixed connections for post bases and also talk about the product design process.
Lateral forces from wind and seismic loads on a structure are typically resisted by a lateral-force-resisting system. There are three main systems used for ordinary rectangular structures: (a) braced frames, (b) moment frames and (c) shearwalls. Moment frames resist lateral forces through bending in the frame members. Moment frames allow for open frames by eliminating the need for vertical bracing or knee bracing. Moment resistance or fixity at the column base is achieved by providing translational and rotational resistance. The new patent-pending Simpson Strong-Tie® MPBZ moment post base is specifically designed to provide moment resistance for columns and posts. An innovative overlapping sleeve design encapsulates the post, helping to resist rotation at its base.
The allowable loads we publish have what I call “triple backup.” This backup consists of Finite Element Analysis (FEA), code-compliant calculations and test data. Here are descriptions of what I mean by that.
Finite Element Analysis Confirmation
Once a preliminary design for the product is developed, FEA is performed to confirm that the product behaves as we expect it to in different load conditions. Several iterations are run to come up with the most efficient design.
Load calculations are prepared in accordance with the latest industry standards. The connector limit states are calculated for the wood-post-to-MPBZ connection and for MPBZ anchorage in concrete. Steel tensile strength is determined in accordance with ICC-ES AC398 and AISI S100-07. Wood connection strength is determined in accordance with ICC-ES AC398 and AC13. Fastener design is analyzed as per NDS. SDS screw values are analyzed using known allowable values per code report ESR-2236. The available moment capacity of the post base fastened to the wood member is calculated in accordance with the applicable bearing capacity of the post and lateral design strength of the fasteners per the NDS or ESR values. Concrete anchorage pull-out strength is determined in accordance with AC398.
Test Data Verification
The moment post base is tested for anchorage in both cracked and uncracked concrete in accordance with ICC-ES AC398.
The moment post base assembly is tested for connection strength in accordance with ICC-ES AC13.
The assembly (post and MPBZ) is tested for various loading conditions: download, uplift and lateral load in both orthographic directions and moment. Applicable factor(s) of safety are applied, and the controlling load for each load condition is published in the Simpson Strong-Tie Wood Construction Connectors Catalog.
Now let’s take a look at a sign post base design example to see how the MPBZ data can be used.
The MPB44Z is used to support a 9ʹ-tall 4×4 post with a 2ʹ x 2ʹ sign mounted at the top. The wind load acting on the surface of the sign is determined to be 100 lb. The MPB44Z is installed into concrete that is assumed to be cracked.
The design lateral load due to wind at the MPB44Z is 100 lb.
The design moment due to wind at the MPB44Z is (100 lb.) x (8 ft.) = 800 ft.-lb.
The Allowable Loads for the MPB44Z are:
Lateral (F1) = 1,280 lb.
Moment (M) = 985 ft.-lb.
Simultaneous Load Check:
800/985 + 100/1,280 = 0.89. This is less than 1.0 and is therefore acceptable.
We are very excited about our new MPBZ! We hope that this product will get you excited about your next open-structure design. Let us know your thoughts by providing comments here.
This week’s post comes from Dan Harmon, an R&D engineer for Simpson Strong-Tie’s Infrastructure-Commercial-Industrial (ICI) group. Dan specializes in post-installed concrete anchor design and spent a decade managing Simpson’s anchor testing lab, where he developed extensive knowledge of anchor behavior and performance. He has a Bachelor of Science in mechanical engineering from the University of Illinois Urbana-Champaign.
Designers and engineers can spend hundreds of hours on detailed drawings of structures, but there are often conditions and coordination that can change well-planned details and drawings. As we all know, paper and reality don’t always agree. Anchorage locations can move as a result of unforeseen circumstances such as encountering reinforcing bars in an existing concrete slab or interference between different utility trades.
With post-installed anchors, one particular jobsite change may require abandoning a hole that has been drilled, leaving the final anchor location adjacent to the abandoned hole. When a hole for an anchor is drilled but never used, it essentially creates a large void in the concrete. Depending on where this void is located in relation to an installed anchor, there is potential for the capacity of that anchor to be reduced. To give guidance on this situation to specifiers, users and contractors, Simpson Strong-Tie conducted a large series of tests in their ISO 17025–accredited Anchor Systems Test Lab in Addison, Illinois.
To evaluate the effect of abandoned holes located adjacent to post-installed anchors, we performed tension tests meeting the requirements of ASTM E488-15 (see Figure 1). A variety of anchor types with common diameters were tested:
Drop-in anchors (1/2″ and 3/4″ diameter)
Wedge-type anchors (1/2″ and 3/4″ diameter)
Concrete screws (1/2″ diameter)
Adhesive anchors with threaded rod (1/2″ diameter)
Each anchor type and diameter was tested under five different conditions:
No abandoned hole near the installed anchor. This is considered the reference condition to which other tests are to be compared.
One abandoned hole at a distance of two times the hole diameter (2d) away from the installed anchor. See Figure 2.
One abandoned hole at a distance of four times the hole diameter (4d) away from the installed anchor.
Two abandoned holes, each at a distance of two times the hole diameter (2d) away from the installed anchor. In test conditions with two holes, the holes were located on either side of the installed anchor, approximately 180º from each other. See Figure 3.
Two abandoned holes, each at a distance of two times the hole diameter (2d) away from the installed anchor, with the holes refilled with a concrete anchoring adhesive that was allowed to cure fully prior to testing. See Figure 4.
This test program is summarized in Table 1. In all cases, the abandoned hole was of the same diameter and depth as the hole prescribed for the installed anchor.
Five tests for each anchor under each condition were tested, and the mean and coefficient of variance of each data set were calculated. These calculated values were used to compare the different conditions.
Across the different anchor types and diameters, the test results showed a number of general rules that held true.
Abandoned holes that are 2” or more away from the anchor have little to no effect on the tension performance of the anchor. Compared to the reference condition with no abandoned hole near the anchor, conditions where the abandoned hole was sufficiently far away were found to be essentially equivalent. This equivalence held true even for anchor types that create expansion forces (drop-in and wedge-type anchors) during their installation.
Two abandoned holes have the same effect on performances as one, regardless of distance from the anchor. This testing showed that adding a second abandoned hole near an installed anchor did not adversely affect tension performance in a significant way. Even within distances of 2 inches, performance did not drop substantially – if at all – in conditions involving two abandoned holes as compared to one.
Filling abandoned holes with an anchoring adhesive prior to installation of the anchor improves performance. In all cases tested, filling abandoned holes with adhesives resulted in increased performance compared to leaving the holes empty. In a majority of cases, performance with filled holes was equivalent to performance in the reference condition regardless of the distance from the anchor.
When the abandoned hole is more than two times the drilled hole diameter but less than 2″from the anchor – and left unfilled – the testing showed a loss in performance. Not surprisingly, the degree of that loss was dependent on the type of anchor. Table 2 shows the capacity reduction compared to the reference condition in testing with expansion anchors. Table 3 shows the same results for concrete screws and adhesive anchors. Conservative suggested performance reductions in these conditions would be 20% for expansion anchors and 10% for concrete screws and adhesive anchors.
In an ideal world, the engineer’s designs could be followed at all times at the jobsite. But we don’t live in an ideal world. Good engineering judgment is needed in situations where variation is required, and having data to support those decisions is always helpful. In the case of abandoned holes near post-installed anchors, it’s Simpson Strong-Tie’s hope that this testing provides additional guidance for the designer, inspector, and jobsite worker.