Welcome to our Structural Engineering Blog! I’m Paul McEntee, Engineering R&D Manager at Simpson Strong-Tie. We’ll cover a variety of structural engineering topics here that I hope interest you and help with your projects and work. Social media is “uncharted territory” for a lot of us (me included!), but we here at Simpson Strong-Tie think this is a good way to connect and even start useful discussions among our peers in a way that’s easy to use and doesn’t take up too much of your time. Continue reading
They say you never forget your first love. Well, I remember my first earthquake, too. My elementary school had earthquake and fire drills often, but the Livermore Earthquake in January, 1980 was the first time we had to drop and cover during an actual earthquake. The earthquake occurred along the Greenville fault and over 20 years later, I was the project engineer for an event center not far from this fault. I don’t think that earthquake that led me on the path to become a structural engineer. I was only seven and was more focused on basketball and Atari games than future fields of study.
My favorite part about the Livermore Earthquake was the 9-day sleepover we managed to negotiate with my parents. I have a big family, so we had a large, sturdy dinner table. My brother Neil and I convinced my parents it would be better if we slept under the table, in case there was an aftershock. And, of course, we should invite our friends, the Stevensons, to sleepover because they don’t have as large a dinner table to sleep under at their house. And it worked! In our defense, there were a lot of aftershocks and an additional earthquake a few days later.
Each year, an earthquake preparedness event known as the Great ShakeOut Earthquake Drill takes place around the globe. The event provides an opportunity for people in homes, schools, businesses and other organizations to practice what to do during earthquakes.
Simpson Strong-Tie is helping increase awareness about earthquake safety and encouraging our customers to participate in the Great ShakeOut, which takes place next Thursday on October 15. It’s the largest earthquake drill in the world. More than 39 million people around the world have already registered on the site.
Earthquake risk is not just a California issue. According to the USGS, structures in 42 of 50 states are at risk for seismic damage. As many of you know, we have done a considerable amount of earthquake research, and are committed to helping our customers build safer, stronger homes and buildings. We continue to conduct extensive testing at our state-of-the-art Tye Gilb lab in Stockton, California, and next Wednesday, we’ll be performing a multi-story wall shake table test for a group of building officials at our lab. We are also working with the City of San Francisco to offer education and retrofit solutions to address their mandatory soft-story building retrofit ordinance and have created a section on our website to give building owners and engineers information to help them meet the requirements of the ordinance.
Our research is often in conjunction with academia. In 2009, we partnered with Colorado State University to help lead the world’s largest earthquake shake table test in Japan, demonstrating that mid-rise wood-frame buildings can be designed and built to withstand major earthquakes.
Earthquake articles like the one from The New Yorker also remind us how important it is to retrofit homes and buildings and to make sure homes, businesses, families and coworkers are prepared.
Like others in our industry, structural engineers play a role in increasing awareness about earthquake safety. We’d like to hear your thoughts about designing and retrofitting buildings to be earthquake resilient. Let us know in the comments below. And if your office hasn’t signed up for the Great ShakeOut Earthquake Drill, we encourage you to do so by visiting shakeout.org.
Vertical deflection resulting from live and dead loads – of both roof and floor framing components – is an important serviceability consideration in the overall design of the building. And while this could be a blog topic in and of itself, this post is instead going to focus on two other types of truss movements that often prompt questions: seasonal up-and-down movement (of the trusses relative to the walls) and horizontal movement (of scissor trusses).
On the one hand, these are completely different topics. But on the other hand, they both deal with movement; which needs to be properly addressed when incorporating trusses into the overall building. So it’s sensible to discuss them together in one blog post.
Seasonal Up-and-Down Movement
This type of movement goes by many different names that might sound familiar – truss arching, truss uplift, partition separation, or – to use the most formal name – ceiling-floor partition separation. All of these names describe the separation that develops between interior partition walls and ceiling finishes, which can cause gaps in the drywall to open in the winter and close in the summer. This movement is often considered to be a truss issue; however, it is not always the trusses that do the moving, but rather the walls or floors, or both, beneath the trusses.
This issue is also not limited to truss construction, but can also occur with other types of wood construction. The truss industry has information on this topic to help educate the market about the causes of ceiling-floor partition separation, best practices and construction techniques for minimizing the movement, and how to accommodate this movement in the structure to prevent drywall cracking.
For those who are interested in a very thorough and technical discussion of this issue and all of the factors that can contribute to it, there is a Technical Note available from the Truss Plate Institute (TPI) called Ceiling-Floor Partition Separation: What Is It and Why Is It Occurring? Although it was written several years ago (by the Small Homes Council-Building Research Council), the information remains relevant because the problem and its causes are the same now as they were then. The Technical Note discusses the potential causes of ceiling-floor partition separation, which may include one or more of the following: attic moisture (and the differential shrinkage and swelling of truss chords due to seasonal changes in moisture content), foundation settlement, expansive soils, excessive cumulative shrinkage of wood framing members and errors made during the construction process such as pulling the camber out of a truss to attach it to a partition. There is even an Appendix with a brief discussion of longitudinal shrinkage and an example calculation showing how much upward deflection results when a truss arches because of differential shrinkage.
For a condensed version, there is also a document available from the Structural Building Components Association (SBCA) called “Partition Separation Prevention and Solutions (How to Minimize Callbacks Due to Gypsum Cracking at the Wall/Ceiling Interface)”. This single-page document is particularly useful for educating the industry to take the appropriate preventive measures during construction, which help minimize problems later.
For example, the use of slotted roof truss clips – such as our STC (see below) – is one preventive measure, since these clips allow for vertical movement, but still provide lateral support at the top of the wall. DS drywall clips can be used in conjunction with the STC clips to secure the drywall to the wall. Then, to allow the drywall ceiling to “float,” the drywall is not fastened to the bottom chord within 16” from the wall. Taking these steps allows movement between the truss and the wall, without causing cracking in the drywall at the wall/ceiling interface.
It is important to note that, while foundation settlement may indicate a structural problem and can be prevented by proper design, truss arching resulting from the natural shrinking/swelling of wood does not indicate any structural problem and cannot be avoided in the design process.
Horizontal Movement of Scissor Trusses
In the typical design of a scissor truss, a pin-type bearing is used at one end, and a roller-type bearing is used at the other end, which results in some amount of horizontal deflection at the roller bearing.
The bearing assumptions used in the design of a scissor truss are important not only to the truss, but they also have design implications for the building as well. Using a pin-type bearing at both ends of the truss has undoubtedly been a temptation to every truss technician at one time or another, when the same scissor truss that is failing the analysis suddenly works as soon as the bearings are switched from pin-roller to pin-pin. Unfortunately, that isn’t a valid option unless the walls are infinitely stiff (which they typically aren’t), or unless special measures are taken to resist the horizontal thrust that develops at the pinned reactions. In most cases, such measures won’t be taken which means with the exception of some rare cases, scissor trusses must be designed with pin-roller bearings.
The horizontal deflection that results when a scissor truss is designed with a roller bearing on one end prompts further questions and discussion. What happens when a scissor truss is rigidly secured to the walls of the building – how does that horizontal movement happen? How much horizontal movement is too much? Should the scissor truss be attached to the wall with a sliding (roller-like) connection?
First, a scissor truss that is rigidly secured to both walls will still experience horizontal movement due to the flexibility of the building’s construction in most residential and light commercial construction. How much horizontal movement is too much for the building? This is definitely a question that the Building Designer needs to answer based on his/her evaluation of the overall structure. However, there are a couple of resources that can provide some insight.
ANSI/TPI 1 has the following provision:
Per ANSI/TPI 1, a scissor truss can have up to 1.25″ of total horizontal deflection in the absence of stricter limits from the Building Designer. Scissor trusses may even be designed with more than this amount of horizontal deflection, along with a warning that special provisions for lateral movement may be required. It is important for the Building Designer to be aware of the calculated horizontal movement of the scissor truss, as reported on the truss design drawing, to ensure that it is an acceptable amount of horizontal movement for the supporting structure and/or to determine whether special provisions for the lateral movement need to be made.
While 1.25″ of total horizontal deflection may seem like a lot of horizontal movement, these calculated horizontal deflections are considered to be conservative; many Designers agree that the predicted movement from the pin-roller bearing combination is greater than will actually occur in the constructed building. This is based on the fact that the design loads may be overstated and the contribution of the sheathing (and drywall if applicable) to resist the horizontal movement is not taken into account during the analysis of the truss.
The National Building Code of Canada (NBC) references Section 5.4.4 of the 2009 Engineering Guide for Wood Frame Construction, which limits lateral movement at the top of each wall to h/500. This correlates to a total allowable horizontal movement of 3/8″ for 8ˈ walls. However, the Canadian truss design standard (TPIC-2014) permits trusses to have a horizontal deflection (at the roller support) of up to 1″. In this case, since the horizontal deflection of the truss exceeds the allowable horizontal deflection of the wall, a sliding connection needs to be used between the truss and the wall.
There are different opinions on the use of sliding connections, such as the slotted TC24 or TC26 connectors (see below), which allow for horizontal movement of the trusses without pushing out the wall, and also provide uplift resistance. The use of these clips also varies greatly by region. There are many places where these clips are used regularly and successfully. However, some Designers prefer to restrict the truss horizontal deflection and require the use of a positive connection between the scissor truss and the wall plate due to concerns regarding the transfer of lateral loads from the top of wall to the roof diaphragm. When TC connectors are used, they are often used on alternating ends of the trusses so that there is a positive connection along each wall at every other truss. Some Designers feel this approach minimizes the horizontal movement between the truss and the wall after the building is constructed and fully sheathed and braced.
There is not a single correct answer to address horizontal truss movement for every building. The amount of horizontal movement that is acceptable for the structure and whether or not a sliding connection should be used will depend on the building, the loading conditions, the designer’s experience and/or judgment, and, in some cases, the local building jurisdiction. What is more important than the decision to either restrict horizontal deflection or utilize sliding connectors like the TC24/TC26 (both have been successful) is that the bearing assumptions used in the design of the scissor truss are accounted for in the design of the building. The worst-case scenario is when a scissor truss is designed with a pin-pin bearing and installed in a building where absolutely no measures have been taken to supply the needed resistance to the calculated horizontal thrust.
What are your thoughts or experiences with either seasonal up-and-down movement or horizontal movement? Let us know in the comments below!
This week’s post comes from Brad Erickson, who is the Engineering Manager for the Composite Strengthening Systems™ product line at our home office. Brad is a licensed civil and structural engineer in the State of California and has worked in the engineering field for more than 17 years. After graduating from Cal Poly, San Luis Obispo with a B.S. in Architectural Engineering, he worked for Watry Design, Inc. as an Associate Principal before coming to Simpson Strong-Tie. Brad is the Engineering Manager for Composite Strengthening Systems and his experience includes FRP design, masonry and both post-tensioned and conventional concrete design. While not at work, Brad enjoys spending time carting his three kids around to their competitive soccer games and practices.
Have you ever had a concrete or masonry design project where rebar was left out of a pour? Chances are, the answer is yes. Did you wish you could solve this problem by putting rebar on the outside of that element? That’s exactly what Simpson Strong-Tie Composite Strengthening Systems™ (CSS) can do for you and your project. In effect, composites act like external rebar for your concrete or masonry element. Composites can be used in similar configurations to rebar but are applied on the exterior surface of the element being strengthened.
The initial offering in our CSS line is our fiber-reinforced polymer (FRP) product group. An FRP composite is created by taking carbon or glass fabric and saturating it with a two-part epoxy which, when cured, creates the composite. Together, the weight of the fabric and the number of layers in the composite determine how much strength it will add to your concrete or masonry element.
Another form of FRP composite is a precured carbon laminate. The carbon fibers are saturated in the manufacturing facility and are attached to the structure using CSS-EP epoxy paste and filler, an epoxy with a peanut butter–like consistency. We also carry paste profilers (pictured below) that help contractors apply the proper amount of paste to a piece of precured laminate.
Of course, before any strengthening project can succeed, proper surface preparation is of the utmost importance. Without a good bond with the substrate, a composite will not be able to achieve the intended performance. Concrete voids must be repaired, cracks must be injected and sealed, and any deteriorated rebar must be cleaned and coated. Prior to composite placement, the surface of the substrate must be prepared to CSP-3 (concrete surface profile) in accordance with ICRI Guideline No. 310.2. Grinding and blasting are the most common surface-preparation techniques.
The following are just a few applications where composites can be used for concrete and/or masonry retrofits. The orange arrows show the direction of the fibers in the fabric – in other words, the direction in which the composite provides tension reinforcement.
This is a summary of the basics of composites and their installation on strengthening projects. As composites are not yet in the design codes in the United States, the American Concrete Institute has produced 440.2R-08: Guide for the Design and Construction of Externally Bonded FRP Systems for Strengthening Concrete Structures. This guide has numerous recommendations for using fiber-reinforced polymer systems to strengthen your concrete or masonry construction.
If you have any questions about composites, post a comment below and I’ll be happy to respond.
Modern code-listed adhesive anchors offer high-strength connection solutions for a variety of applications. However, as in all construction projects, good product performance requires proper selection and installation. In this blog post, we will discuss the challenge of installation orientation and an accessory that can help installers more easily make proper adhesive anchor installations—the piston plug adhesive delivery system.
ACI 318-11 Appendix D (Anchoring to Concrete) calculations use a uniform bond stress model to calculate an adhesive anchor’s resistance to bond failure. According to this theory, an adhesive anchor is assumed to transfer applied loads into the concrete base material uniformly along its effective embedment depth, hef. The equation for an anchor’s basic bond strength (expressed in pounds of force) is simply the adhesive formulation’s bond strength per unit area (λ * τcr) multiplied by the idealized cylindrical surface area of the insert that is in contact with the adhesive (π * da * hef):
Nba = λ τcr π da hef (ACI 318-11, Eq. D-22)
Although the model is a simplification of reality, the mathematical expression represents the core assumption that the adhesive is able to transfer stress completely along the entire depth of the anchorage. This is a key requirement in installation: Anchoring adhesives must be installed such that air entrapment and significant voids are prevented.
Downward installations (Figure 1) have historically presented relatively few challenges for adhesive injection in this regard. In such applications, gravity is helpful; the adhesive naturally flows to the bottom of the drilled hole while being dispensed from the cartridge through a static mixing nozzle. The installer maintains the open end of the nozzle below the free surface of the adhesive until the drilled hole is filled to the desired level. For deep holes, extension tubing is affixed to the open end of the nozzle to increase reach. This procedure avoids entrapping air bubbles in the adhesive material.
Installations into horizontal, upwardly inclined or overhead drilled holes (Figure 2) require more care on the part of adhesive anchor installers. Although the installation principle to avoid entrapping air is similar for these orientations, a key difference is that gravity does not help to keep the adhesive towards the “bottom” (deepest point) of the drilled hole. At worst, it can work against the installer when ambient temperatures may cause the adhesive to run out of the hole during injection. These adhesive anchor installations can be more difficult for an untrained installer and can slow the rate of work. This is one of the reasons that ACI 318-11 Section D.9.2.4 requires continuous special inspection of adhesive anchor installations in these three orientations when the application is also intended to resist sustained loads.
To aid the installer, Simpson Strong-Tie offers a piston plug adhesive delivery system (Figure 3). Consisting of pre-packaged flexible tubing, piston plugs and an adhesive retaining cap, this system allows installers to more easily and consistently make high-quality installations while completing their work efficiently. The installation sequence is provided in Figure 4.
The system consists of three components:
- Piston plug – The key component of the system, it is slightly smaller in diameter than the drilled hole. As the adhesive is dispensed into the drilled hole, the piston plug is displaced out of the hole by the advancing volume of the injected adhesive. The displacement creates a more positive feel for the installer to know where the free surface of the adhesive is.
- Flexible tubing – For use with the piston plug to facilitate injection at the deepest point of the drilled hole.
- Adhesive retaining cap – Provided to prevent adhesive material from flowing out of the drilled hole after dispensing and to provide a centering mechanism for the insert. For heavy inserts in overhead conditions, other means must be provided to carry the weight of the insert and prevent it from falling or becoming dislodged from the hole before the adhesive has fully cured.
What do you think about the piston plug adhesive delivery system? Let us know by posting a comment below.
If you’re one of the many engineers still confused by the ACI 318 – 11 Appendix D design provisions, this blog will help explain what’s required to achieve a ductile performing anchorage. Most building codes currently reference ACI 318 – 11 Appendix D as the required provision for designing a wide variety of anchor types that include expansion, undercut, adhesive and cast-in-place anchors in concrete base materials. This blog post will focus on section D.220.127.116.11(a) for an anchor located in a high seismic region. We’ll go over what these requirements are with a simple design example.
Ductility is a benefit in seismic design. A ductile anchor system is one that exhibits a meaningful degree of deformation before failure occurs. However, ductility is distinct from an equally important dimension called strength. Add strength, and a ductile steel element like the one shown in Figure 1 can now exhibit toughness. During a serious earthquake, a structural system with appreciable toughness (i.e., one that possesses both strength and ductility in sufficient degree) can be expected to absorb a tremendous amount of energy as the material plastically deforms and increases the likelihood that an outright failure won’t occur. Any visible deformations could help determine if repair is necessary.
Let’s start off with a simple example that will cover the essential requirements for achieving ductility and applies to any type of structural anchor used in concrete. We’ll arbitrarily choose a post-installed adhesive anchor. This type of anchor is very common in concrete construction and is used for making structural and nonstructural connections that include anchorage of sill plates and holdowns for shear walls, equipment, racks, architectural/mechanical/electrical components and, very frequently, rebar dowels for making section enlargements. We’ll assume the anchor is limited to resisting earthquake loading in tension only and is in seismic design category C – F. Section D.18.104.22.168 requires that if the strength-level earthquake force exceeds 20% of the total factored load, that the anchor be designed in accordance with section D.22.214.171.124 and D.126.96.36.199. We will focus on achieving the ductility option, (a), of D.188.8.131.52.
To understand anchor ductility we need to first identify the possible failure modes of an anchor. Figure 2 shows the three types of failure modes we can expect for an adhesive anchor located away from a free edge. These three failure modes generically apply to virtually any type of anchor (expansion, screw, cast-in-place or undercut). Breakout (Nb) and pullout (Na) are not considered ductile failure modes. Breakout failure (Nb) can occur very suddenly and behaves mostly linear elastic and consequently absorbs a relatively small amount of energy. After pullout failure (Na) has been initiated, the load/displacement behavior of the anchor can be unpredictable, and furthermore, no reliable mechanism exists for plastic deformation to take place. So we’re left with steel (Nsa). To achieve ductility, not only does the steel need to be made of a ductile material but the steel must govern out of the three failure modes. Additionally, the anchor system must be designed so that steel failure governs by a comfortable margin. Breakout and pullout can never control while the steel yields and plastically deforms. This is what is meant by meeting the ductility requirements of Appendix D.
Getting back to our design example, we have a single post-installed 5/8” diameter ASTM F1554 Gr. 36 threaded rod that’s embedded 12” deep, in a dry hole, in a concrete element that has a compressive strength of 2,500 psi. The concrete is 18” thick and we assume that the edge distance is large enough to be irrelevant. For this size anchor, the published characteristic bond strength is 743 psi. Anchor software calculations will produce the following information:
The governing design strength is compared to a demand or load combination that’s defined elsewhere in the code.
Here’s the question: Before proceeding with the remainder of this blog, judging by the design strength values shown above, should we consider this anchorage ductile? Your intuition might tell you that it’s not ductile. Why? Pullout clearly governs (i.e., steel does not). So it might come as a surprise to learn that this adhesive anchor actually is ductile!
To understand why, we need to look at the nominal strength (not the design strength) of the different anchor failure modes. But first let’s examine the equations used to determine the design strength values above:
The above values incorporate the notation φ (“phi”) and a mandatory 0.75 reduction factor for nonductile failure modes (Ncb ,Na) for applications located in high seismic areas (seismic design category C–F). The φ factor is defined in section D.4. However, manufacturers will list factors specific to their adhesive based on anchor testing. The mandatory 0.75 reduction comes from section D.184.108.40.206 and is meant to account for any reduction associated with concrete damage during earthquake loading. The important thing to remember is that the nominal strength provides a better representation of the relative capacity of the different failure modes. Remove these reduction factors and we get the following:
Now steel governs since it has the lowest strength. But we’re not done yet. Section D.220.127.116.11.(a).1 of Appendix D requires that the expected steel strength be used in design when checking for ductility. This is done by increasing the specified steel strength by 20%. This is to account for the fact that F1554 Gr. 36 threaded rod, for example, will probably have an ultimate tensile strength greater than the specified 58,000 psi. (Interestingly, the ultimate strength of the ½” threaded rod tested in Figure 1 is roughly 74 ksi, which is about 27% greater than 58,000 psi.) With this in mind, the next step would be to additionally meet section D.18.104.22.168.(a).2 such that the following is met:
By increasing the steel strength by 20%, the nominal strength of the nonductile failure modes (Ncb ,Na) must be at least that much greater to help ensure that a ductile anchor system can be achieved. The values to compare finally become:
Now steel governs, but one more thing is required. As shown in Figure 3, Section D.22.214.171.124.(a).3 of Appendix D also requires that the rod be made of ductile steel and have a stretch length of at least eight times the insert diameter (8d). Appendix D defines a ductile steel element as exhibiting an elongation of at least 14% and a reduction in area of at least 30%. ASTM F1554 meets this requirement for all three grades of steel (Grade 36, 55 and 105) with the exception of Grade 55 for anchor nominal sizes greater than 2”. Research has shown that a sufficient stretch length helps ensure that an anchor can experience significant yielding and plastic deformation during tensile loading. The threaded rod shown in Figure 1 was tested using a stretch length of 4” (8d). Lastly, section D.126.96.36.199.(a).4 requires that the anchor be engineered to protect against buckling.
Appendix D doesn’t require that an anchor system behave ductilely. Three additional options exist for Designers in section D188.8.131.52. Option (b) allows for the design of an alternate failure mechanism that behaves ductilely. Designing a base plate (or support) that plastically hinges to exhibit ductile performance is one example. Option (c) involves a case where there’s a limit to how much load can be delivered to the anchor. Although option (c) under D.184.108.40.206 falls under the tensile loading section of Appendix D, the best example would apply to anchorage used to secure a wood sill plate or cold-formed steel track. We know from experiments that the wood crushes or the steel yields and locally buckles at a force less than the capacity of the concrete anchorage. Clearly energy is absorbed in the process. The most commonly used option is (d), which amplifies the earthquake load by Ωo. Ωo can be found in ASCE 7 – 10 for both structural and nonstructural components. The value of Ωo is typically taken to be equal to 2.5 (2.0 for storage racks) and is intended to make the anchor system behave linear elastically for the expected design-level earthquake demand.
These same options exist for shear loading cases. However, achieving system ductility through anchor steel is no longer an option for shear loading according to ACI 318 – 11, because the material probably won’t deform appreciably enough to be considered ductile.
While factors such as edge-distance and embedment-depth restrictions make achieving ductility difficult for post-installed anchors, it should come as some consolation that in many cases the Designer can achieve ductile performance for cast-in-place anchors loaded in tension through creative detailing of reinforcing steel (section D.5.2.9) to eliminate breakout as a possible failure mode. This has been explored in some detail in two previous Simpson Strong-Tie blogs titled “Anchor Reinforcement for Concrete Podium Slabs” and “Steel Strong Wall Footings Just Got a Little Slimmer.”
What are your thoughts? Visit the blog and leave a comment!
Designing built-up columns? Now there’s a way to mechanically laminate multiple 2x members to meet the specifications in the NDS. Simpson Strong-Tie evaluated Strong-Drive® SDW Truss-Ply screws for attaching multiple laminations with easier installation methods. With these screws, there’s no longer a need to nail from both sides of the column, or to use not-so-common 30d nails as specified in the NDS, or to pre-drill for bolts. Instead, installers can now install all the screws from one side of the built-up column, which provides time and cost savings.
Columns can be classified into solid columns, built-up columns and spaced columns. Solid columns are single members or individual members glued together to act as one solid member. Mechanically laminated built-up columns are formed by fastening two or more members with bolts, nails or screws. If built-up members have spacer blocks between the members, they create a spaced column. The design of built-up columns is different from the design of solid columns.
These three classes of columns have differing load capacities. The capacity of a built-up column can be expressed as a percentage of the strength of a solid column of the same dimensions and made with material of the same grade and species. The ratio of the built-up column compression capacity to that of a solid column is defined as efficiency (1). The efficiency of built-up columns is 1.0 in the strong axis and between 60 and 75 percent in the weak axis depending on the type of fastening. The loss in capacity in the weak axis compared to a solid column is due to the slip between the laminations.
Column failure is due to crushing, buckling, or a combination of both modes. Short columns more often experience crushing failure, and long columns tend to fail more often by buckling. According to the NDS, the efficiencies are generally higher in short and long columns than in intermediate columns. The NDS assigns nailed built-up intermediate columns a 60% efficiency and bolted built-up intermediate columns a 75% efficiency. Even though long and short columns would have higher efficiencies, all column lengths are assigned a single efficiency. Note that provisions in NDS 220.127.116.11 allow short columns to use full design values when designed as individual columns.
Whole-section engineered wood products are recommended for higher compression loads, although they can add cost.
Design of solid columns is addressed in Section 3.7 of the NDS. The main difference between solid column and built-up column capacity is in the calculation of Cp, the column stability factor. The column stability factor adjusts for the buckling effect on the column capacity. If the column is completely braced in all directions, then Cp can be taken as 1. For all other conditions, Cp should to be evaluated for both strong-axis and weak-axis bracing conditions. In solid columns, the column stability factor is calculated as follows:
In this calculation, le/d represents the larger of the ratios l1/d1 and l2/d2 as shown in Figure 1. The slenderness ratio of solid columns, le/d, shall not exceed 50. Higher slenderness ratios have a lower Cp factor, which means that a slender column can buckle more easily and has lower compression capacity than a similar column with a lower slenderness ratio. The same holds true for built-up columns.
Built-up columns fastened with nails or bolts are addressed in Sections 15.3.3 and 15.3.4 of the NDS. However, fastening built-up column members with screws is not addressed in the NDS. For built-up columns, the only difference in design compared to solid columns is the addition of Kf, a column stability coefficient, in the calculation of Cp. See Figure 2 for built-up column notation. For built-up columns, Cp is calculated as follows:
The Cp value is calculated for slenderness ratios based on l1/d1 and l2/d2, and the smaller Cp is used to calculate the adjusted compression design value parallel to grain. In the strong axis, Kf = 1, and the design is similar to solid columns. However, in the weak axis buckling is affected by the slip and load transfer that occurs at fasteners between the laminations, and the Kf factor changes with the type of fastener.
NDS Section 15.3.1 provides the limitations for built-up columns based on these design attributes:
- Each lamination has a rectangular cross section and is at least 1-1/2” thick,
- All laminations are of same depth and faces of adjacent laminations are in contact,
- All laminations are full column length.
These limitations apply to laminations fastened with nails and bolts. In Simpson Strong-Tie design method they also apply to Strong-Drive SDW screws.
The spacing and end distance requirements for nails are covered in Section 18.104.22.168 of the NDS. The nails need to be driven from opposite sides of the column and need to penetrate at least ¾ of the thickness of last lamination. If all of the requirements are met, Kf of 0.6 can be used in the calculation of Cp, when l2/d2 is the limiting ratio for calculation of FcE. The NDS does not provide a table for built-up column capacities fastened with nails. The designer has to run through calculations and follow the provisions of NDS Section 15.3.3 to determine the capacity of a nailed built-up column.
Let’s figure out the nail length needed to connect 3 – 2xmembers. For a 3-ply member the nail length needs to be a minimum of 2 x 1.5inches +3/4 x 1.5 inches = 4.125 inches. Only 30d or higher nails are available in these lengths. Since these nails are not commonly used in the job site and do not fit the regular nail gun, installers may need to use a special nail gun. The NDS provides some typical nailing schedules that are shown here in Figure 3.
NDS Section 22.214.171.124 has end, edge distance and spacing requirements for bolts. Also a metal plate or washer is required between the wood and the bolt head and between the wood and the nut. The nuts should be tightened to ensure that the faces of adjacent laminations are in contact. Figure 4 is a detail showing the typical bolting schedules. If the requirements of the NDS Section 15.3.4 are met, Kf of 0.75 can be used in the calculation of Cp, when l2/d2 is the limiting ratio for calculation of FcE. The NDS does not provide built-up column capacities fastened with bolts. Again the designer has to determine these capacities by calculation.
New Option – Fastening with Simpson Strong-Tie® Strong-Drive® TRUSS –PLY SDW Screws
Simpson Strong-Tie® tested column assemblies as shown in the Figure 5 to determine Column Stability Coefficient, Kf, for SDW screws. The limitations of NDS Section 15.3.1 have been found to also apply to Strong-Drive Truss-Ply SDW screws. The spacing and end distance requirements for the screws are shown in Figure 5. One huge advantage of using SDW screws is the screws can be installed from one side of the column or from both sides of the column. Installation from one side or both sides affects the Kf factor used in the calculation of Cp. If the screws are installed from one side of the column, then Kf of 0.6 can be used in the calculation of Cp when l2/d2 is used to calculate FcE. If the screws are installed from both sides of the column, then Kf of 0.7 can be used in the calculation of Cp when l2/d2 is the limiting ratio for calculation of FcE.
Let’s work on a design example for built-up columns fastened with Strong-Drive® Truss-Ply® SDW screws:
Example: Calculate the capacity of a 3-2×6 built-up member attached with SDW screws with a) installation of screws from only one side b) Installation of screws from both sides of the column.
Column Type: Built-up
Column Length: 10 ft.
Bracing: Completely unbraced in both directions
Size if Column: 3 – 2 x 6
Wood Species: SPF
Load Duration Factor, CD: 1
Temperature Factor, Ct: 1
Wet Service Factor, CM: 1
Per Table 4.3.1 (table shown below) of NDS:
Fc’ = Fc x CD x CM x Ct x CF x Cp
Per Table 4A or 4B of NDS, Compression parallel to grain, Fc = 1150 psi
Emin = 510000 psi
Size Factor CF = 1.1
Fc* = Fc x CD x CM x Ct x CF 1265 psi
Now let’s calculate Cp in both directions:
Cp in Strong-Axis –
Fc* = reference compression design value parallel-to-grain multiplied by all applicable modification factors except Cp (see 2.3 of NDS)
FcE = 0.822 Emin‘/ (l1/d1)2
l1 = 120 in
d1 = 5.5 in
FcE = 881 psi
Kf = 1.0 for solid columns and for built-up columns where l1/d1 is used to calculate FcE and the built-up columns are either nailed or bolted
c = 0.8 for sawn lumber
Substituting the values above, Cp = 0.557
Cp in Weak-Axis –
Fc* = reference compression design value parallel-to-grain multiplied by all applicable modification factors except Cp (see 2.3 of NDS)
FcE = 0.822 Emin‘/ (l2/d2)2
l2 = 120 in
d2 = 4.5 in
FcE = 590 psi
Kf = 0.6 for built-up columns fastened with SDW screws from one side of column
Kf = 0.7 for built-up columns fastened with SDW screws from both sides of column
c = 0.8 for sawn lumber
Substituting the values above with Kf = 0.6, Cp = 0.246 (Screws installed from one side)
Substituting the values above with Kf = 0.7, Cp = 0.287 (Screws installed from both sides)
For screws installed from same side, minimum Cp = Minimum (0.557, 0.246) = 0.246
Column Capacity = Fc* x Cp x d1 x d2= 7690 lbs.
For screws installed from both sides, minimum Cp = Minimum (0.557, 0.287) = 0.287
Column Capacity = Fc* x Cp x d1 x d2= 8970 lbs.
To avoid these long calculation steps and to help the designer, Simpson Strong-Tie compiled a table with allowable compression capacities for built-up columns made with several typical combinations of No. 2 visually graded lumber and fastened with SDW screws. Now that you have a new and faster way of fastening multiple plies using SDW screws along with an easy to use design table, go ahead and design away!
If you have any questions or comments about fastening built-up columns with Simpson Strong-Tie fasteners, pass them along to us in the Engineering Department.
- Malhotra, S.K and A.P Sukumar, A Simplified Procedure for Built-up Wood Compression members, St. John’s, Newfoundland, Annual Conference, Canadian Society for Civil Engineering, June 1-18, 1989.
- Malhotra, S.K and D.B. Van Dyer, Rational Approach to the Design of Built-Up Timber Columns, Madison, WI, Forest Products Research Society (Forest Products Society), Wood Science, Vol. 9, No. 4: 174-186, 1977.
- National Design Specification for Wood Construction (NDS), ANSI/AWC NDS-2012. 2012. American Wood Council, Leesburg, VA. 282 pp
This week is the 10th anniversary of Hurricane Katrina, and we have all seen articles on the lessons learned from the storm. Engineers learn something new from every storm. However, I think that Hurricane Katrina just gave us some very strong reminders of things we already knew.
Hurricane Katrina reminded us that hurricanes are flood events as well as high-wind events. And I don’t mean the flooding in New Orleans. No, I mean the flooding along the Gulf Coast from Louisiana to Florida.
I witnessed the complete devastation of the Mississippi Gulf Coast from Waveland to Biloxi. Structures within the first few (and often many) blocks from the beach were simply flattened by water. Fortunately, these areas are coming back, but the structures being built there now bear little resemblance to the homes that graced the beach 10 years ago.
I remember my father-in-law having his new house built on the coast in Waveland more than 20 years ago. As a young engineer, I gave it the once over and noted that the builder had connected the roof framing to the top plate, but little else. I made some recommendations, such as continuing the connections down throughout the rest of the house to the foundation. The builder followed my suggestions and then presented my father-in-law with the bill “for your son-in-law the inspector.” He was happy to pay it. Nevertheless, although the house was wind resistant, it could not stand up to the rushing waters from Hurricane Katrina.
Katrina reminds us that the only way to get away from floods, other than not building near the water, is to elevate structures above them. Due to flood regulations, new houses along the Gulf Coast are now elevated high in the air, in the hope of avoiding flooding from future storms. Simpson Strong-Tie is proud to have developed some products during the last few years that make it easier to build structures elevated on pilings.
One such product is our CCQM column cap that strengthens the connection of support beams to masonry piers. Another is the Strong-Drive® SDWH Timber-Hex HDG structural screw, which is meant to replace through-bolts to make the connection of a beam to a wood piling easier and more reliable.
Hurricane Katrina reminds us of the value of building codes. After the storm, the LSU Hurricane Center conducted a number of simulation studies on the effect of a direct, Katrina-like storm on the states of Louisiana, Mississippi and Alabama. The simulations were run on the existing stock of buildings, and then run again on the same stock of buildings, assuming that certain features that result from modern building codes were present. These features included shutters or impact-resistant windows, enhanced nailing of the roof deck to the roof framing, framing connected together with hurricane clips and straps to achieve a continuous load path. In addition, in the Louisiana study, a secondary water barrier over the joints in the roof sheathing was added.
The studies found that the decrease in wind damage from the simulated storms was astounding. In Louisiana, the study showed a 79% reduction in economic losses due to wind. In Alabama, the study revealed a 72% reduction in economic losses due to wind. The Gulf states seem to have received the message loud and clear. In the years following Hurricane Katrina, Louisiana adopted a statewide building code and Mississippi adopted a uniform building code for the four counties along the coast. Recently, Alabama has also adopted a statewide residential and energy code. But in general, building codes are still quite varied in coastal states. This report from the Insurance Institute for Business and Home Safety evaluates the effectiveness of building codes in coastal states.
Finally, Hurricane Katrina reminds those of us who do damage surveys that you need to know what you are getting into before you go. As soon as the storm hit and we saw the scope of the damage, four members of the Simpson Strong-Tie Engineering Department in our McKinney, Texas, office decided we needed to go see the damage first-hand before any repairs were made. So two days after the storm struck, off we went to Jackson, Mississippi. There, we rented two vans stocked up with food, water and fuel. Unfortunately, the fuel and the food/water ended up in separate vans. Before long, we were separated in traffic and could not communicate due to loss of cell signal.
Our team spent two days viewing the damage first-hand along the Louisiana and Mississippi coast, but spent a lot of time our last day trying to find some fuel so we could make it back to Jackson. I remember spending the night in a hotel without power full of storm victims, and then months later receiving the bill and being charged for a movie!
What do you remember from Hurricane Katrina? Let us know in the comments below.
Have you ever been at home during an earthquake and the lights turned off due to a loss of power? Imagine what it would be like to be in a hospital on an operating table during an earthquake or for a ceiling to fall on you while you are lying on your hospital bed.
One of the last things you want is to experience serious electrical, mechanical or plumbing failures during or after a seismic event. During the 1994 Northridge earthquake, 80%-90% of the damage to buildings was to nonstructural components. Ten key hospitals in the area were temporarily inoperable primarily because of water damage, broken glass, dangling light fixtures or lack of emergency power.
ASCE 7 has an entire chapter titled Seismic Design Requirement of Nonstructural Components (Chapter 13 of ASCE 7-10) that is devoted to provisions on seismic bracing of nonstructural components. Unfortunately, not a lot of Designers are aware of this part of the ASCE. This blog post will walk Designers through the ASCE 7 requirements.
Nonstructural components consist of architectural, mechanical, electrical and plumbing utilities. Chapter 13 of ASCE 7-10 establishes the minimum design criteria for nonstructural components permanently attached to structures. First, we need to introduce some of the terminology that is used in Chapter 13 of ASCE 7.
- Component – the mechanical equipment or utility.
- Support – the method to transfer the loads from the component to the structure.
- Attachment – the method of actual attachment to the structure.
- Importance Factor (Ip) – identifies which components are required to be fully functioning during and after a seismic event. This factor also identifies components that may contain toxic chemicals, explosive substances, or hazardous material in excess of certain quantities. This is typically determined by the Designer.
Section 13.2.1 of ASCE 7 requires architectural, mechanical and electrical components to be designed and anchored per criteria listed in Table 13.2-1 below.
Architectural components consist of furniture, interior partition walls, ceilings, lights, fans, exterior cladding, exterior walls, etc. This list may seem minor compared to structural components, but if these components are not properly secured, they can fall and hurt the occupants or prevent them from escaping a building during a seismic event. The risk of fire also increases during an earthquake, further endangering the occupants.
Section 13.5 of ASCE 7-10 includes the necessary requirements for seismic bracing of architectural components. Table 13.5-1 provides various architectural components and the seismic coefficients required to determine the force level the attachments and supports are to be designed for.
Mechanical and electrical components consist of floor-mounted and suspended equipment. It also includes suspended distributed utilities such as ducts, pipes or conduits. These components are essential in providing the necessary functions of a building. In a hospital, these components are required to be fully functioning both during and after a seismic event. A disruption of these components can make an entire hospital building unusable. In order for hospitals to properly service the needs of the public after a seismic event, fully functioning equipment is essential.
Section 13.6 of ASCE 7-10 provides the requirements of seismic bracing for mechanical and electrical components. Table 13.6-1 provides a list of typical components and the coefficients required to determine the force level the attachments and supports are to be designed for.
Chapter 13 lists some typical requirements for which components are to be anchored and supported under specific conditions:
- Section 13.1.4 item 6c: Any component weighing more than 400 pounds.
- Section 13.1.4 item 6c: Any component where its center of gravity is more than 4 feet above the floor.
- Section 126.96.36.199 has specific electrical conduit size and weight requirements.
- Section 13.6.7 has specific size and weight requirements for suspended duct systems.
- Section 13.6.7 has specific size and weight requirements for suspended piping systems.
The chapter also has some general exceptions to the rules:
- 12 Inch Rule: When a distributed system such as conduit ducts or pipes are suspended from the structure with hangers less than 12 inches in length, seismic bracing is not required.
- If the support carrying multiple pipes or conduits weighs less than 10 pound/feet of lineal weight of the component, the seismic bracing of the support does not have to be considered.
These exceptions do have limitations that are clearly listed in Sections 188.8.131.52, 13.6.7 and 13.6.8.
These systems may not seem important in the structural systems of a building, but they are essential in allowing the building to function the way it was designed to serve the public. It is also important that occupants are able to escape a damaged building after a seismic event. Obstacles such as bookcases blocking exit doors or falling debris may prevent occupants from leaving a building after a seismic event.
It is important that Designers are aware of these code requirements and take the time to read and understand what is needed to provide a safe structure.
A deck and porch study reported that 33% of deck failure-related injuries over the 5-year study period were attributed to guard or railing failures. While the importance of a deck guard is widely known, there was a significant omission from my May 2014 post on Wood-framed Deck Design Resources for Engineers regarding the design of deck guards.
A good starting point for information about wood-framed guard posts is a two-part article published in the October 2014 and January 2015 issues of Civil + Structural Engineer magazine. “Building Strong Guards, Part 1” provides an overview of typical wood-framed decks, the related code requirements and several examples that aim to demonstrate code-compliance through an analysis approach. The article discusses the difficulties in making an adequate connection at the bottom of a guard post, which involve countering the moment generated by the live load being applied at the top of the post. Other connections in a typical guard are not as difficult to design through analysis. This is due to common component geometries resulting in the rails and balusters/in-fill being simple-supported rather than cantilevered. “Building Strong Guards, Part 2” provides information on the testing approach to demonstrate code-compliance. Information about code requirements and testing criteria are included in the article as well.
Research and commentary from Virginia Tech on the performance of several tested guard post details for residential applications (36” guard height above decking) is featured in an article titled “Tested Guardrail Post Connections for Residential Decks” in the July 2007 issue of Structure magazine. Research showed that the common construction practice of attaching a 4×4 guard post to a 2x band joist with either ½” diameter lag screws or bolts, fell significantly below the 500 pound horizontal load target due to inadequate load transfer from the band joist into the surrounding deck floor framing. Ultimately, the research found that anchoring the post with a holdown installed horizontally provided enough leverage to meet the target load. The article also discussed the importance of testing to 500 pounds (which provides a safety factor of 2.5 over the 200-pound code live load), and the testing with a horizontal outward load to represent the worst-case safety scenario of a person falling away from the deck surface.
Simpson Strong-Tie has tested several connection options for a guard post at the typical 36” height, subjected to a horizontal outward load. Holdown solutions are included in our T-GRDRLPST10 technical bulletin. In response to recent industry interest, guard post details utilizing blocking and Strong-Drive® SDWS TIMBER screws have been developed (see picture below for a test view) and recently released in the engineering letter L-F-SDWSGRD15. The number of screws and the blocking shown are a reflection of the issue previously identified by the Virginia Tech researchers – an adequate load path must be provided to have sufficient support.
Have you found any other resources that have been helpful in your guard post designs? Let us know by posting a comment.
Simpson Strong-Tie is sponsoring the 24th Short Course on Cold-Formed Steel Structures hosted by the Wei-Wen Yu Center for Cold-Formed Steel Structures (CCFSS). The course will be held on October 27-29, 2015 at the Drury Plaza Hotel at the Arch in St. Louis, MO.
This three-day course is for engineers who have limited or no experience designing with cold-formed steel (CFS), as well as those with experience who would like to expand their knowledge of cold-formed steel structural design. Lectures will be given by industry-recognized experts Roger LaBoube, Ph.D., P.E., and Sutton Stephens, Ph.D., P.E., S.E. The course is based on the 2012 AISI North American Specification for the Design of Cold-Formed Steel Structural Members and the 2012 North American Standards for Cold-Formed Steel Framing. Dr. Wei-Wen Yu’s book Cold-Formed Steel Design (4th Edition) will be a reference text.
The course will address such topics as design of wall studs, floor joists, purlins, girts, decks and panels. It is eligible for 2.4 Continuing Education Units (CEUs). Advance registration is requested by October 10, 2015. For more information and to register, click here.